SEMINAR REPORT ON EVALUATION OF FLOW LIQUEFACTION AND LIQUEFIED SHEAR STRENGTH
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SEMINAR REPORT ON EVALUATION OF FLOW LIQUEFACTION AND LIQUEFIED SHEAR STRENGTH USING CONE PENETRATION TEST
SEMINAR REPORT ON EVALUATION OF FLOW LIQUEFACTION AND LIQUEFIED SHEAR STRENGTH USING CONE PENETRATION TEST.doc (Size: 264 KB / Downloads: 148)
Flow liquefaction is a major design issue for large soil structures such as mine tailings impoundments and earth dams. If a soil is strain softening in undrained shear and, hence, susceptible to flow liquefaction, an estimate of the resulting liquefied shear strength is required for stability analyses. Many procedures have been published for estimating the residual or liquefied shear strength of cohesionless soils. This paper presents cone penetration test-based relationships to evaluate the susceptibility to strength loss and liquefied shear strength for a wide range of soils. Liquefaction of soil is a phenomenon by which the strength and stiffness of a soil is reduced by earth quake shaking or other rapid loading.
Soil liquefaction is a major concern for structures constructed with or on saturated sandy soils. The phenomenon of soil liquefaction has been recognized for many years. Since 1964, much work has been carried out to explain and understand soil liquefaction. Much of the work in the past three decades has been on liquefaction induced during earthquake loading i.e., cyclic liquefaction.
Robertson and Wride distinguished between liquefaction due to cyclic loading, where the effective overburden stress can reach zero during cyclic loading with a resulting loss of soil stiffness (cyclic liquefaction/softening), and liquefaction due to strain softening with a resulting loss of shear strength (flow liquefaction) and presented a simplified flowchart to aid in the evaluation. Flow liquefaction is also referred to as static liquefaction.Flow liquefaction can occur in any saturated or near saturated metastable soil such as very loose cohesionless sands and silts as well as very sensitive clays. For failure of a soil structure, such as a slope or embankment, a sufficient volume of material must strain soften. The resulting failure can be a slide or a flow depending on the material characteristics of the soils and ground geometry.The resulting movements are due to internal, gravity-induced stresses and can occur after the triggering mechanism occurs.
Flow liquefaction is a major design issue for large soil structures such as mine tailings impoundments and earth dams. A liquefaction analysis of sloping ground i.e., ground subject to a static driving shear stress is a challenge for geotechnical engineers. Many procedures have been published for estimating the residual or liquefied shear strength of cohesionless soils. However, procedures based on case histories remain the most popular Olson and Stark suggested a detailed procedure, based on an extensive database of case histories, consisting of three tasks
1 evaluate susceptibility to strength loss
2 evaluate liquefied shear strength, suliq, and postliquefaction
3 evaluate if liquefaction will be triggered.
Olson and Stark defined the liquefied shear strength, suliq, as the shear strength mobilized at large deformation by a saturated contractive soil following the triggering of a strain-softening response. Others have used the term undrained residual shear strength. The term liquefied shear strength will be used in this paper to be consistent with the more recent Olson and Stark terminology. The Olson and Stark procedure uses normalized penetration resistance with no correction for soil type.
Case histories have shown that when significant strength loss occurs in critical sections of a soil structure, failures are often rapid, occur with little warning, and the resulting deformations are often very large. Experience has also shown that the trigger events can be very small. Triggering should always be assumed if the soils are susceptible to strength loss. Hence, the design for high risk soil structures should be carried out with caution. In general the emphasis in design is primarily on the evaluation of susceptibility to strength loss and the resulting liquefied shear strength.
This paper presents cone penetration test (CPT)-based relationships to evaluate the susceptibility to strength loss and liquefied shear strength for a wide range of soils. Case-history analyses by a number of investigators are reviewed and used with some additional newer case histories. Emphasis is placed on more recent case histories that had modern CPT measurements available.Extrapolations beyond the case-history data are guided by laboratory studies and theory.
This section deals with normalized cone penetration test parameters.Robertson developed a chart to identify soil behavior type SBT based on normalized CPT parameters as shown in Fig. 1. The CPT parameters are normalized by the effective overburden stress to produce dimensionless parameters, Qt and Fr, where
Qt = (qt - σvo)σ’vo
Fr = [ fs(qt -σvo)]100%
where qt = CPT corrected total cone resistance
fs = CPT sleeve friction
σvo = preinsertion in situ total vertical stress and
σ’vo = preinsertion in situ effective vertical stress.
SBT index, Ic, could represent the SBT zones in the Qt - Fr chart where Ic is essentially the radius of concentric circles that define the boundaries of soil type. Robertson and Wride modified the definition of Ic to apply to the Robertson Qt - Fr chart as defined by
Ic = [ (3.47 − log Qt)2 + (log Fr + 1.22)2]0.5
Robertson and Wride suggested a normalized cone parameter to evaluate soil liquefaction, using normalization with a variable stress exponent, n, where
Qtn = [(qt –σvo)pa]( paσ’vo)n
where (qt -σvo) pa = dimensionless net cone resistance;
(paσ’vo)n = stress normalization factor
n = stress exponent that varies with SBT and
pa = atmospheric pressure in the same units as qt and σv.
When n =1, Qtn = Qt. the stress exponent, n, could be estimated using the SBT index, Ic, and that Ic should be defined using Qtn. Contours of Ic are included in Fig. 1 to illustrate the trend.
In recent years there have been several publications regarding the appropriate stress . All the methods agree that in the clean sand region of the Qtn-Fr SBT chart _Zone 6_ the stress exponent is typically close to 0.5 and in the clay region of the SBT chart _Zone 3_ the stress exponent is close to 1.0.
Robertson suggested that the above stress exponent would capture the correct in situ state for soils at high stress level and that this would also avoid any additional stress level correction for liquefaction analyses. It is well recognized that the normalized cone resistance decreases as a soil becomes more fine grained due to the increasing compressibility of fine-grained soils compared to coarse-grained soils. This was identified by Robertson where the normally consolidated region on the CPT SBT chart extends down the chart; i.e., as soil becomes more fine grained the normalized cone resistance Qtn decreases and Fr increases. Robertson and Wride suggested that the soil behavior index Ic increases when soils become more fine grained and that when Ic>2.60 soils tend to be more claylike. Independent studies have confirmed that most samples are claylike when Ic>2.60.
Robertson updated the trends in normalized cone parameters with overconsolidation ratio (OCR), sensitivity (St), and age as shown in Fig. 1. In fine-grained soils, the normalized cone resistance Qtn increases with increasing OCR with little influence on the normalized friction ratio (Fr). On the other hand, Fr decreases with increasing St, with little influence on Qtn. Both Qtn and Fr tend to increase as soils become stiffer and stronger with age.
Olson and Stark presented a comprehensive summary of case histories where flow liquefaction occurred.Most of the case histories presented by Olson and Stark did not have detailed CPT records that included sleeve friction measurements. However, there are six 6 case histories where flow liquefaction occurred and where detailed electric CPT records are available. The three additional new case histories are Jamuna Bridge, Sullivan Tailings, and Canadian Mine.
CANADIAN MINE DESIGN
During the preliminary design stage for a proposed open-pit mine in Canada, a trial excavation was made to evaluate the stability of the natural surface soils. The surface soils in one region were primarily composed of soft, low plastic, sensitive silty clay extending to a depth of about 10–15 m. The silty clay had an average plasticity index with a high liquidity index. The high LI is consistent with the high sensitivity and low residual/liquefied shear strength. During the excavation, a 50-m-wide flow failure occurred within the sensitive clay in a 5-m-high, 50-degree slope. The upper 2.5 m of soil was frozen. The flow slide was retrogressive in nature and the failed soil moved into the excavation along an almost horizontal surface. The size of failed mass almost doubled in size over a 4-day period.
The six case history sites where CPT measurements were available including sleeve friction values are identified as Class A, since they are the most reliable in terms of CPT data. Some older case histories had either mechanical CPT (MCPT) or electric CPT but no friction sleeve values and are identified as Class B. Class B results are less reliable than Class A records in terms of the CPT data. Other case histories used values estimated by Olson from either SPT, relative density, or best estimates and are identified as either Class C, Class D, or Class E, respectively. Classes C, D, and E are not reliable in terms of CPT results, since no CPT measurements were made.
Liquefaction is a phenomenon in which the strength and stiffness of a soil is reduced by earthquake shaking or other rapid loading. Liquefaction and related phenomena have been responsible for tremendous amounts of damage in historical earthquakes around the world. Liquefaction occurs in saturated soils, that is, soils in which the space between individual particles is completely filled with water. This water exerts a pressure on the soil particles that influences how tightly the particles themselves are pressed together. The term liquefaction has actually been used to describe a number of related phenomena. Because the phenomena can have similar effects, it can be difficult to distinguish between them. The mechanisms causing them, however, are different. These phenomena can be divided into two main categories: flow liquefaction and cyclic mobility.
Cyclic mobility is a liquefaction phenomenon, triggered by cyclic loading, occuring in soil deposits with static shear stresses lower than the soil strength. Deformations due to cyclic mobility develop incrementally because of static and dynamic stresses that exist during an earthquake. Lateral spreading, a common result of cyclic mobility, can occur on gently sloping and on flat ground close to rivers and lakes. The 1976 Gautemala earthquakecaused lateral spreading along the Motagua river
On level ground, the high porewater pressure caused by liquefaction can cause porewater to flow rapidly to the ground surface. This flow can occur both during and after an earthquake. If the flowing porewater rises quickly enough, it can carry sand particles through cracks up to the surface, where they are deposited in the form of sand volcanoes or sand boils.
Flow liquefaction is a phenomenon in which the static equilibrium is destroyed by static or dynamic loads in a soil deposit with low residual strength. Residual strength is the strength of a liquefied soil. Static loading, for example, can be applied by new buildings on a slope that exert additional forces on the soil beneath the foundations. Earthquakes, blasting, and pile driving are all example of dynamic loads that could trigger flow liquefaction. Once triggered, the strength of a soil susceptible to flow liquefaction is no longer sufficient to withstand the static stresses that were acting on the soil before the disturbance.
An analogy can be seen in the picture above, where the static stability of a ski jumper in the starting gate is disturbed when the jumper pushes himself from the start seat. After this relatively small disturbance, the static driving force caused by gravity, being greater than the frictional resisting force between the ski and snow, causes the skier to accelerate down the ramp. The path that brings the ski jumper to an unstable state is analogous to the static or dynamic disturbance that triggers flow liquefaction - in both cases, a relatively small disturbance proceeds an instability that allows gravity to take over and produce large rapid movements.
Failures caused by flow liquefaction are often characterized by large and rapid movements which can produce the type of disastrous effects experienced by the Kawagishi-cho apartment buildings, which suffered a remarkable bearing capacity failure during the Niigata earth quake . The Turnagain Heights landslide;Alaska earth quake which is thought to be triggered by liquefaction of sand lenses provides another example of flow liquefaction.
As these cases illustrate, flow failures can involve the flow of considerable volumes of material, which undergoes very large movements that are actually driven by static stresses. The disturbance needed to trigger flow liquefaction can, in some instances be small.
EVALUATING FLOW LIQUEFACTION
Since flow liquefaction requires a strain-softening soil response and strength loss, the evaluation of susceptibility to flow liquefaction is controlled by an evaluation of the potential for a soil to strain soften in undrained shear. Experience has shown that very loose sands and very sensitive low-PI clays can experience an abrupt loss of strength at small shear strains resulting in low undrained shear strength. Although many natural high-PI clays can also experience some strength loss, they tend to be more ductile and experience more gradual loss of strength at larger shear strains. The key element to identify a soil’s susceptibility to flow liquefaction is to identify very loose coarsegrained soils i.e., sand, silty sands, and sandy silts and very sensitive fine-grained soils like silts, silty clays, clayey silts, and clays.
The concepts for strength loss and liquefied shear strength in sands stem from the work on the critical void ratio by Casagrande . Castro expanded the basic concept of a critical void ratio and developed the term steady state to define the liquefied strength. At about the same time, the concepts of critical state soil mechanics were under development. Both concepts recognize that the state of a soil is represented by a combination of the void ratio and effective stress and that if a soil is loose of either steady or critical state the soil can strain soften in undrained shear.
Critical state soil mechanics was used to develop the state parameter (Ψ) concept and applied these concepts to the CPT results and to soil liquefaction . The state parameter (Ψ) is defined as the difference between the in situ void ratio, eo, and the void ratio at critical state, ecs, at the same mean effective stress, p’.Jefferies and Been provided a detailed description of the evaluation of the soil state using the CPT and show that the inverse problem of evaluating the state from the CPT response is complex and depends on several soil parameters.For high risk project and implimentations, a detailed interpretation of the CPT results using laboratory results and numerical modeling can be appropriate, although soil variability can complicate the interpretation procedure.
For low risk project and implimentations and in the initial screening for high risk project and implimentations, it is often appropriate to estimate the state of the soil directly from the CPT results. A screening method was developed to estimate the state of the soil using the normalized CPT results. In a general sense, soils with a state denser than the critical state (Ψ<0) will be dilative and will strain harden in undrained shear, whereas soils with a state looser than the critical state (Ψ>0) will be contractive and will strain soften in undrained shear. When a soil has a state parameter Ψ>-.05, strain softening and strength loss in undrained shear can be expected. Hence, defining a region on the CPT SBT chart that represents a state parameter of about -.05 is helpful as a screening technique to identify the susceptibility for flow liquefaction
DIALATIVE AND CONTRACTIVE SOIL RESPONSE
Combined with the test results from frozen samples Robertson , it is possible to identify a zone on the normalized SBT, based on Qtn and Fr, that represents the approximate boundary between the dilative and contractive soil response as shown in Fig. 3. Also identified in Fig. 3 is the region that defines normally consolidated clays with a sensitivity of 1 or more based on the work of Robertson . Robertson showed that as the soil sensitivity increases in fine-grained soils, the measured CPT normalized friction ratio Fr decreases as indicated in Fig. 3. There is a clear trend for the normalized cone resistance to decrease as Ic increases for the same soil state.
Fig. 3 indicates that the region in the lower left portion of the SBT chart defines soils that are likely susceptible to contractive behavior and strength loss in undrained shear. Included in Fig. 3 is the approximate boundary between the contractive and dilative soil response based on the normalized penetration resistance. This criteria vary slightly with effective overburden stress and a range is presented. It is clear from Fig. 3 that the criterion suggested by Olson and Stark applies only to clean sands where typically Fr<1%.
Robertson and Wride suggested a correction factor to correct the normalized cone resistance in silty sands to an equivalentclean sand value (Qtn,cs) using the following
Qtn,cs = KcQtn
where Kc = a correction factor that is a function of grain characteristics of the soil that can be estimated using Ic
CONTOURS OF EUIVALENT CLEAN SAND NORMALISED CONE RESISTANCE
Fig. 4 shows the contours of the equivalent clean sand cone resistance, Qtn,cs, on the CPT SBT chart. The contours of Qtn,cs follow a trend similar to the dilative-contractive boundary defined in Fig. 3 and that a value of Qtn,cs between 50 and 70 likely represents the boundary between the contractive and dilative state for a wide range of soils. Robertson indicated that the contours of Qtn,cs are essentially contours of the state parameter (Ψ). Hence, soils with a constant value of Qtn,cs have essentially a similar state parameter and hence a similar response to loading.
LIQUEFIED SHEAR STRENGTH
The concept that plasticity index of soils can be defined as a range of water contents producing variation in undrained shear strength has been experimentally verified with the help of a large number of tests on soils of diverse nature. This has led to the redefinition of the plastic limit as the water content at which undrained shear strength is around 170 kN/m2. Undrained shear strength of a soil at the liquid limit can be considered to be around 1.7 kN/m2. Accordingly, both the liquid limit and the plastic limit have been determined in the present work by a single consistent method. The undrained shear strength-water content relationship has been found to be log-linear for a wide range of water contents beginning from lower than the plastic limit to higher than the liquid limit. This resulted in the formulation of an expression for predicting undrained shear strength of a remolded soil at any water content based solely on its liquid limit and plastic limit.
A critical parameter in the evaluation of the liquefaction of soils is the residual or liquefied shear strength. This liquefied shear strength determines the magnitude of the deformation that the soil will undergo once it has liquefied. Current procedures for estimating the liquefied shear strength are based on laboratory testing programs, or from the back-analysis of case histories of liquefaction failures where in-situ test data were available. However, it has several limitations including the very limited amount of data available, the significant uncertainties involved in the back-calculation of the liquefied shear strengths, and the lack of consistent and rational methods in the use of the available data.
EVALUATION OF LIQUEFIED SHEAR STRENGTH
Estimating liquefied shear strength values suliq from failure case histories uses stability calculations that require many simplifying assumptions and idealizations. Most approaches use limit equilibrium methods based on either prefailure or postfailure geometry and may or may not include inertial effects. Many case histories involved retrogressive sliding that is rarely accounted for in the back-analyses. Hence, the evaluation of the liquefied shear strength based on case histories is often approximate at best. There is a theoretical link between the state parameter and the liquefied undrained shear strength ratio suliq σvo. Since clean sand equivalent normalized penetration resistance Qtn,cs is essentially equivalent to the state parameter, values of Qtn,cs are compared to suliq σvo for the case histories.
When kinetics is considered, the back-calculated shear strength values were larger than if kinetics were not included. A weighted average prefailure vertical effective stress is used to estimate the liquefied shear strength. Previous studies have presented back-calculated values of the liquefied shear strength ratios to three decimal places. This level of perceived accuracy is inconsistent with the many uncertainties in the backcalculation of flow slides and this study has shown “best estimated” values to only two decimal places. For consistency the best estimate values those computed by Olson and Stark but rounded to two decimal places.
When soils are strain hardening in undrained shear, the undrained shear strength will typically exceed the drained shear strength, although cavitation can limit the full value of the undrained strength. The strength ratio in terms of drained shear strength (in simple shear) can be represented as
τ/σ’vo = tanφ’
The drained strength ratio in triaxial compression can be larger than defined by this equation. Hence, for soils that are strain hardening in undrained shear, i.e., where Qtn,cs>70, a conservative low estimate of the undrained strength ratio is about 0.4–0.5.
Fig. 7 shows the best estimate values for the liquefied strength ratios and mean CPT clean sand equivalent penetration resistance values for Classes A and B case histories. The more reliable CPT results from the case histories that are Class A are shown in large solid symbols and Class B by the large shaded symbols. Based on the observation that no case history, with reliable measured CPT results, had a mean clean sand equivalent normalized penetration resistance, Qtn,cs>70 and the observation that most soils are strain hardening (i.e., not susceptible to strength loss)in undrained shear when Qtn,cs>70 with τ σvo≈0.4.
Based on the above observations, the proposed relationship trends toward suliq σvo =0.4 at Qtn,cs=70.
A conservative, essentially lower bound, relationship was selected to capture much of the variability in the back-calculated liquefied shear strength values from the case histories and to recognize the need for caution in soils where Qtn,cs<70.
The proposed relationship can be conservatively low in sensitive clays, where the remolded shear strength ratio su® σ’vo can be defined by
su® σ’vo = fsσ’vo = (FrQtn)100
It is based on the observation that the remolded shear strength for most clay soils is approximately equal to the CPT sleeve friction, fs (Robertson) .When the proposed relationship is applied to all CPT results within a sounding, it is recommended that the average value of liquefied shear strength within a noninterlayered soil deposit that is considered to be susceptible to strength loss should be applied for stability analyses. Hence, the variability within one soil type is captured.
A lower bound relationship between the liquefied shear strength and clean sand equivalent normalized penetration resistance is proposed that avoids the need to extrapolate beyond the case-history database. In cases where less permeable layers may inhibit pore pressure dissipation and where void redistribution could occur, a more conservative estimate may be appropriate, and for high risk project and implimentations, more detailed field and analytical studies should be carried out. When the proposed relationship is applied to all CPT results within a noninterlayered soil deposit, it is recommended that the average value of liquefied shear strength should be applied for stability analyses, since the relationship was based on average CPT values within noninterlayered deposits from the case histories.
• An update of the CPT-based criteria is useful to evaluate the susceptibility to flow liquefaction and the liquefied strength .
• The case histories indicate that very young, very loose, nonplastic or low-plastic soils tend to be more susceptible to significant and rapid strength loss than older, denser and more plastic soils.
• A lower bound relationship between the liquefied shear strength and clean sand equivalent normalized penetration resistance is proposed that avoids the need of further studies for moderate project and implimentations.
• When the proposed relationship is applied to all CPT results within a non interlayered soil deposit, it is recommended that the average value of liquefied shear strength should be applied for stability analyses.
• When significant strength loss occurs in critical sections of a soil structure, failures are often rapid, occur with little warning, and the resulting deformations are often very large.
• The CPT is a useful in situ test that can provide continuous estimates of the potential for flow liquefaction.
• CPT-based approach is a simplified method that should be used appropriately depending on the risk of the project and implimentation.
• For low risk project and implimentations, the CPT-based method is appropriate when combined with selective samples to confirm the soil type as well as conservative estimates of soil response.
• For moderate risk project and implimentations, the CPT-based method should be combined with appropriate additional in situ testing, as well as selected undisturbed sampling and laboratory testing, to confirm soil response.
• For high risk project and implimentations, the CPT-based method should be used as an initial screening to identify the extent and nature of potential problems followed by additional in situ testing and appropriate laboratory testing on high-quality samples.
Robertson, P. K. _1990_. “Soil classification using the cone penetration
test.” Can. Geotech. J., 27_1_, 151–158.
Robertson, P. K. _2009a_. “Discussion of ‘CPT-based probabilistic soil
characterization and classification’ by K. Onder Cetin and Cem
Ozan.” J. Geotech. Geoenviron. Eng., 135_1_, 84–107.
Robertson, P. K. _2009b_. “Interpretation of cone penetration tests—A
unified approach.” Can. Geotech. J., 46, 1337–1355.
CANADIAN MINE DESIGN
EVALUATING FLOW LIQUIEFACTION
DIALATIVE AND CONTRACTIVE SOIL RESPONSE
CONTOURS OF EQUIVALENT CLEAN SAND
NORMALISED CONE RESISTANCE
LIQUEFIED SHEAR STRENGTH
EVALUATION OF LIQUEFIED SHEAR STRENGTH
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